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About ReinisGailitis

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  • Birthday 09/25/89

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    Riga, Latvia
  1. The particular one you posted is just a commentary and does not contain a pdf file or similar. There are some AD's that are written in this way and are posted only as written info, others are pdf. All AD's are quite short, 1 to possibly 3 pages. Cheers
  2. In the lower part of each of the descriptions there is a link to open a pdf file with full info.
  3. Try different browser. Also, have you tried just the ?
  4. I'll repeat myself - go and download them, they're free to everyone - Cheers
  5. Go to and download the yourself. Cheers
  6. From my knowledge I would say yes, if you pour the slab in two pours and allow long enough time between them the lower part of the slab should shrink a considerable rate leaving only a small residual shrinkage strain to take place after second pour is constructed. One must consider the fact that the actual thickness of the second pour will be more that 250mm, as the lower part of the slab, which was poured to 250mm thickness now is thinner because of shrinkage. Also one should consider the residual shrinkage strain to take place after second part of the slab is constructed. I suppose that if you could show by simple analysis on realistic assumptions that in this situation the shrinkage to which the top steel of the second pour is subjected is equal or just a little higher than one which would prevail in a single pour slab of 250mm, than you may assume good bond conditions. I would think that the margin for this assumption would have to be small, the usual 5% for example, as there are too many factors not really known in design, which will influence the was concrete cures etc. Although if you have to wait a certain time for the first part of the slab to cure before concreting the second you could start having problems with shrinkage strain induced tension from restraint on the top surface of the slab from first pour which would require you to provide some nominal reinforcement, which I suppose would mitigate any economic benefit from using smaller lap lengths in the top steel of the second part of the slab which is poured in the second stage, because if you don't provide reinforcement to control cracking, you may find the slab cracks substantially and possibly reduced slabs resistance to shear force, which may pose a significant threat if there are no links provided. Anyway this is all just from my theoretical understanding of the problem and the only way to know and really "show" that all of this is true is by testing. Cheers
  7. Yes, it is mainly due to bleeding and shrinkage. There are few book that mention the bond conditions but in general there is no such guidance that you seek. Anyway I wonder what is the problem? Can't you find a solution which will satisfy anchorage requirements if bad bond conditions are assumed? P.S. As with everything in Eurocode you would definitely have to test to show that the good conditions actually prevail in the situation.
  8. Good information - thanks!
  9. These equations must be satisfied if you don't want to make an explicit check of the rotational capacity of a particular section (or rather a region of a structure). In general there indeed is no direct restrictions on x/d if one doesn't assume any redistribution of elastic moments in the structure. In any case it would be good practice to design section with ductile behavior, which in general can be accomplished by limit x/d to that given of the equations 5.10 (thus implicitly taking care of rotational capacity of section), because the limit in equations 5.10 will still depend on the amount of moments redistributed. ISE manual also suggests limiting x/d to 0.45 to achieve ductile behavior. Here are my two cents... The thing is that you could design section to be balanced x/d=0,617, though I always remind myself that I'm just a human and human errors do happen in design offices - seen them, made them. So in any case if you design ductile sections there is added safety to the structure, for examples, because if you underestimated settlements of some foundations there may be a situation where hogging bending moments in the support region in a slab are higher in real life than those you've assumed during design, and if that region won't be ductile enough it could lead to failure of the structure. By that I want to say that I always design liming x/d<=0,45 or smaller, if appropriate. Cheers
  10. I'm talking about equations 5.10a and 5.10b - if you rearrange them as to get value of <=x/d, you'll see that it is influenced by two things - redistribution factor and ultimate compression strain of concrete in bending. In general case when delta (redistribution factor) = 1, thus, there is no redistribution of design bending moments, and concrete grade <=C50/55, thus fck<=50MPa, you'll get a value of x/d<=0,448, to be exact - this ensures the ductility of the section. I'm not sure about your spreadsheet, as I don't have time to go through it, but I can tell you that in ULS bending resistance calculation you don't take the resistance of reinforcing bars higher than the yield strength, even if the strain for section analysis is much higher than yield strain. Also, if for reinforcing steel design you use the stress-strain relationship with inclined top branch, which approximates the effect of strain-hardening, then for the region after yield strain you'll have a different value of elastic modulus of reinforcing steel, thus, your analysis of stress-strain in previous post is incorrect, as past yield steel won't have such high elastic modulus and thus the stress won't be as high as you predicted. Cheers
  11. If foundation pads are narrow enough and if there are no extension of pads at the ends of the walls, as well as if the length of walls, under which pads are made are long enough, punching shear won't be the governing failure mode. In certain cases punching shear could be a problem and should be checked, though. An example could be a strip foundation with really wide pads which also extend at the ends of the walls (fox example a free standing length of a wall in the middle of a building) - in such a case the could be a potential punching shear problem either on the end of the wall or around the wall itself (if it is a short wall). Best of luck!
  12. If you're talking about flexural design at ULS, then yes, you don't need to limit the strain of reinforcing steel if you're using design yield strength, because you're limiting the section capacity to that provided by tension force which corresponds with the yield of the reinforcing steel. As long as x/d<=0,45 (in EC for B500 steel and concrete class up to C50/55 actually x/d<=0,617 - this corresponds to a balanced section where steel yields at the same time as concrete crushes) the steel will yield prior to concrete crushing in compression, thus achieving plastic "failure", thus steel will yield prior to concrete reaching its failure strain, and, of course after that there will be large deformations and steel will start to strain-harden to provide some additional resistance until either concrete will crush in compression or steel will reach it's ultimate tensile stress, after which it'll fail. P.S. dik - in EC requirement of x/d<=0,45 (sometimes even 0,30, if large redistribution of bending moments has been assumed) is equivalent to the 0,75*Pbal requirement in North America. Hope this explanation helps!
  13. I'm have not designed seismic buildings till now so I'm not sure what are the code requirements exactly, but it seems to me obvious that if the code requires to amplify all the internal forces and moments in the structures under consideration then of course you would do the same for a connection, as you design it to carry loads in support of the member which is subjected to the combination of the previously mentioned amplified internal forces and moments. If on the other hand code requires you to amply only moment and shear force, for example, leaving axial force as it is (without amplification), than of course you would use the same concept to design connection. Anyway you have to apply engineering judgement to all practical problems you encounter daily. If we talk about the particular problem than of course it seems quite obvious that if you consider a member with a set of internal forces and moments in it, which all are possibly amplified and thus resemble a probably real situation in a real structure, then of course you need to make sure that the support of the particular member (including the connection itself) is able to transfer these internal forces and moments to the foundation, because if it cant, than you may encounter a structural failure of the particular member in the connection. Hope I've helped, best of luck to you.
  14. You need to use the same internal forces and moments in connection design that you used in member design, thus, if you designed member taking into account the amplification because you believe it is necessary or design code requires it, than of course you need to apply the same principle for connection design, as you would not want you design be governed by connection (at least I don't accept such solutions for my designs in general). Cheers
  15. Hallo. In general I check deflection of the floor or roof as a whole, because I'm interested in limiting the total slope of the floor in all possible directions.